Shotcrete Worker Certification

A week or so ago, during a Precon meeting, I was asked by the City of Seattle if I cared if some of the shotcrete nozzle men or Blow-pipe men were trainees.  Without knowing any code requirements on the subject, I said that I did not care as long as they did it right.

It was interesting to me that the City of Seattle asked and recorded the names of all shotcrete workers.  Now as I review the shotcrete shop drawings, I see that all nozzle men and Blow-pipe men are listed along with years of experience and state they meet or exceed the ACI Qualification requrements.  Apparently there is a certification we can request: https://www.shotcrete.org/ASAcertification.htm

Does anyone have further knowledge on the subject of certification or on the design considerations and limitations of using shotcrete applications in lieu of pouring?

SFRS Special Inspections for One- or Two- Family Dwellings

We should be specifying in our construction documents what items need special inspection (SI).  I have seen an itemized list in our general notes (for large projects...not Single Family though) that covers special inspection for general construction (through IBC 1704).  However, we are required per IBC 1705.3 to also include the Seismic-force-resisting system (SFRS).  If we are designing a detached one-or-two family dwelling, we can avoid the SI requirement if our SFRS does not contain certain horizontal or vertical irregularities (listed in Exception#3 of 1705.3).  If it does have any of these irregularities, the SFRS SI is triggered.

Here is the perplexing requirement worth blogging and coming to a consensus on: 

Section 1705.3 requires SI of the "seismic-force-resisting systems..." (Section 1705.31).  

This boggles Chuck and me.   If one was able to clearly delineate out the SFRS in the overall Bearing Wall System, only a few sticks would be left, in my opinion, that are used for gravity only and, subsequently, need not be checked.  We would be asking the special inspector to spend 2 days on the house checking every nail that resides as a component of the SFRS, for shearwalls, diaphragms, etc.

Chuck suggested that perhaps only the exceptions triggering the SI of the SFRS (along with any other important components) be specified for SI.  That seems logical to me.  Perhaps we could substantiate this preference or "interpretation" by alluding to the fact that the phrase "seismic-force-resisting systems" from section 1705.31 is plural and only certain seismic force resisting systems, or "components", require inspection.  

Thoughts?

Concrete Gravity Columns

So I was looking through the ACI 318-08 at the tie requirements for concrete gravity columns and saw that the entire Chapter 21 was re-written.  So don't go looking for revision bars along the side of the pages because there aren't any.  Some changes have been made to the tie requirements that I thought would be nice to summarize.  I'm going to pull information straight from the code so please forgive if it is too redundant.  Thanks

Concrete gravity columns by definition are not part of the lateral force resisting system and therefore are detailed per section 21.13.

Depending on the design method and level of forces, columns are to be details per 21.13.3 or 21.13.4.  If one determines the design displacement, the resulting member and shear forces on the column, and if the capacity is greater than the demand then 21.13.3 can be followed.  Otherwise 21.13.4 must be followed which provides a level of ductility and strength that match a column which is part of the LFRS.

21.13.3 - where induced moments/shears from the design displacements are less than the member moment/shear strength capacity then to following sections shall be satisfied.

• 21.13.3.1 - for lightly loaded members where Pu < Ag f'c/ 10 shall satisfy
• 21.5.2.1 - provides minimum reinforcement for flexural members. 
• Stirrups shall be placed no more then d/2 for length of member.
• 21.13.3.2 - for moderately loaded members where Pu > Ag f'c/10 shall satisfy
• 21.6.3.1 - longitudinal reinforcement to be between 1 and 6 percent.
• 21.6.4.2 - arrangement of hoops and crossties in column, 14" max.
• 21.6.5    - shear strength requirements at faces of joint for column.
• Tie spacing So shall not exceed smaller of 6 times diameter of smallest longitudinal bar and 6 inches.
• 21.13.3.3 - for heavily loaded members where Pu > .35Po shall satisfy
• 21.13.3.2 - see above for moderately loaded column.
• 21.6.4.7 -  cover limitation of 4" along with additional transverse reinforcement to confine additional cover.
• Amount of transverse reinforcement shall be 1/2 that required by 21.6.4.4 but shall be spaced not greater than So for the full member length.

 21.13.4 - where induced moments/shears are not calculated or where the demand exceeds the capacity of the column then the following sections shall apply.  Note that this would be our typical design approach.

• 21.13.4.1 - materials shall satisfy
• 21.1.4.2 - compressive strength f'c shall not be less than 3000 psi.
• 21.1.4.3 - compressive strength for light weight shall not be greater than 5000 psi.
• 21.1.5.2 - deformed reinforcement shall be ASTM A706.
• 21.1.5.5 - value of fy or fyt shall not exceed 60 ksi. 
• Mechanical splices shall satisfy 21.1.6 and welded splices shall satisfy 21.1.7.1
• 21.13.4.2 - for column load Pu < Ag f'c/10 shall satisfy
• 21.5.2.1 - provides minimum reinforcement for flexural members.
• 21.5.4 - shear strength requirements for flexural member
• Stirrups shall be placed no more than d/2 for length of member.
• 21.13.4.3 -  Where gravity axial load exceeds Ag f'c/10 the following shall be satisfied.
• 21.6.3.1 - longitudinal reinforcement to be between 1 and 6 percent.
• 21.6.3.2 - lap splices are permitted only within the center half of the member length, shall be designed as tension lap splices, shall be enclosed transverse reinforcement conforming to 21.6.4.2 and 21.6.4.3.  This is a major change from previous codes. 
• 21.6.4 - all sections must be followed therefore tie reinforcement similar to column that is part of the LFRS.
• 21.6.5 - all sections must be followed therefore shear strength requirements similar to column that is part of the LFRS.
• 21.7.3.1 - joint reinforcement shall satisfy
• 21.6.4.4(a) or 21.6.4.4(b)
• 21.6.4.2
• 21.6.4.3
• 21.6.4.7
• except as permitted by 21.7.3.2 

As can tell from section 21.13.4.3 when the design displacement forces are not checked the detailing requirements become much more significant.

Wood Shearwalls on Podium Slab

Recently received a new comment from Seattle DPD plans reviewer Roussi Roussev. 

"The transfer slab at level 2 supports discontinued wood shear walls, and therefore is required to comply with ASCE 7.12.3.3.3.  Please provide verification of the slab for the maximum force (uplift or compression) that can develop in accordance with the seismic load combinations with overstrength factor."

Were the heck did this come from?  We've been doing two-stage analysis for years on this type of structure and have not had to consider wood shearwalls as discontinuous.  If we look back to the 2003 IBC section 1620.1.2 modified ASCE 7 and specifically exempted light-framed walls supported on concrete slabs from this type of irregularity.  Then the 2006 IBC saw most of the seismic design cut from the code and referenced ASCE 7 without modifications.

My initial response to Roussi was that since a two-stage analysis (ASCE 7 12.2.3.1) was used, only the amplified forces from the upper portion need to be designed for and that no irregularity exists.   This response was not well received.

Charlie spoke with Roussi concerning this comment and he was not really sympathetic to our way of thinking.  In fact Roussi mentioned that both John Siu and Steve Pfeiffer are taking a more rigid definition of shearwalls supported by concrete slabs and that they are indeed an irregularity that needs to be designed for.  He stated that there has been a project where the reinforcement did increase due to the seismic loading. 

In my second response to DPD I modified the PT slab model to include load combinations for seismic.  Below is a list of steps taken to modify the model.

• Added loads "Seismic Loading N-S" and "Seismic Loading E-W".
• Added the four load combinations:
• "Factored LC (N-S):  (1.2 + .2Sds)D + Omega QE +L + .2S"

• "Factored LC (E-W): (1.2 + .2Sds)D + Omega QE +L + .2S"
• "Factored LC (N-S):  (.9 - .2Sds)D + Omega QE

• "Factored LC (E-W): (.9 - .2Sds)D + Omega QE
• The seismic factor of 4.65 accounts for Omega*Rw/Rc*LRFD/ASD.  For all four of the load combinations there is an Alt. Envelope Factor.  Use the same factors except the seismic should be negative.  This has the effect of reversing the seismic load direction.
• Then for each N-S and E-W directions the holdown loads were applied at each end of the shearwall.  Sign convention used was that North and East are compression (down) loads, South and West are tension (up) loads.  Loads used are the allowable tension load from the threaded rod holdown schedule on sheet S1.4.

N-S Shearwalls



E-W Shearwalls

Modifications finished!  Run the model!

Reinforcement requirements did NOT change due to included seismic load combinations.

Provide response to comment along with PT calculations back to DPD.  If we let DPD know that the seismic load combinations don't control then maybe one day they will stop asking for it.

Snow Loading in King County.

So i have reason to believe that the snow loading around King County area may have - or - may be changing...

I was speaking to an engineer for a roofing system that we have designesd in the Bellevue area, and she was suggesting that the 25psf snow white paper did not apply to our project or jurisdiction.

I went online, and was able to download a snow load map that has the Isolines on it ( King County Ground Snow Load Analysis - updated Feb 2005)... ground snow load is determined by multiplying the elevation of the site by the (interpolated) isoline. In my particular case, a quick spot check revealed that my ground snow loads were heading up close to 45psf! that starts getting up past the 25psf roof loads in a hurry.

The King County maps can be found here...(G:\Codes\Snow Loads)

If anyone has any additional knowledge or experience with snow loads around the Puget Sound Area - please reply to this post...

Wood Stud Wall Design - 5psf

Just took a brief poll in the office to talk about the 5psf horizontal loading required by the code for wood stud design on walls per [IBC 1607.13]. Interesting...

The commentray for the IBC 2003 talks about the loading being related to pressurization, or nominal impact loads. The transient nature of these loads seems to indicate the use of the Cd=1.6 factor.

The poll reveals that most of us are using a Cd = 1.6 for our combinations involving wind and seismic as applies to the Fb bending criteria. For the D+L load case, we are NOT applying the 5psf or 1.6 factor.

Steve points out that the 1997 UBC [1611.5] is actually a LOT more clear about the 5psf: It specifically mentions that the 5psf need not be applied simultaneously with the wind or seismic loads, and further gives deflection criteria...

Steve and I speculate that the 5spf may be used in lieu of actual horizontal seimic loads on wall (unless there is a heavy "special case" phenomenon) , and wind should be applied as per standard wind calcs. The question is whether to apply the 5psf as a live load to the D+L load case. When you "factor" (excuse the pun here) in the 97 UBC, it seems that we should consider the 5psf in this case as well (what Cd to use though?)

So to summarize:

1. With the Seismic combination - we are generally using 5spf in lieu of hor seismic loads, with Cd = 1.6 (for flexure and axial)
2. With wind combinations - the wind does not generally apply with the wind loads to a 2x4 wall (the main point of concern)
3. With dead and live combos - we have NOT been applying the 5psf - but should. On this note, we're floating out the idea that we may apply a Cd =1.6 for flexural component of the design only. Cd for other aspects will remain at 1.0.

This seems like a reasonable approach. Let me know what you think.

Hand Rail Design - factors?

The topic of handrail design came up yesterday, and I did a quick pole to find out what folks are up to on this front.

Interesting...

The design forces for the railing are typically indicated in the code - ASCE 7-05 [4.4.1] talks about the loads, and some exceptions for SFR.

In IBC 2006 and previous, [1607.7.1.3] used to talk about a 1/3 stress increase for members and their attachments wrt (with respect to) handrail design. In IBC 2009, this has been removed - drat! (i'm not sure that we were all taking advantage of this stress increase when we had it).

So the question came up about whether the load for handrails referred to in the code was indeed a "live" load, or more of an "impact" load? For wood design, Cd for impact is 2.0, while for live, it is 1.0. Since "common sense" would indicate that handrails loads are not likely to a sustained loading condition, it could be argued that a stress increase could be appropriate - and may be necessary - in order to come up with a reasonable solution.

Maybe we should consider getting the commentary on this section to see why they removed the section.

Any comments?