Headed Hex-Head bolts or J/L bolts - thats the question?

Arbitrarily a discussion came about on this issue.

I was surprised by it, and did not particularly think that there were any significant and "strong" thoughts out there regarding this issue.

I was wrong.

The brief poll i took in the office showed that some folks are using the hex head anchor bolts in place/lieu of J or L bolts at all anchor bolt locations (including shearwalls). The thought offered was related to a reduction of J/L bolts per Appendix D.

App D DOES mention J/L bolts, but only in the context of PULLOUT (as far as i can tell).

The basic thought there is that the pullout of the J/L bolt is related only to the projected area of the "hook" portion, and this "area" is multiplied (arbitrarily) by a 0.9 factor. In contrast the hex head bolt pullout is based on a net "brg" area multiplied by a 8.0 factor. This is apparently based on testing results.

Fair enough.

The main issue here is that the J/L bolts are being spec'd in non-pullout uses such as in sole plate applications to shearwalls - NOT for holdowns - which are themselves again subject to the "nuanced" intricacies of the "infamous" Appendix D.

Contractors appear to not "give a hoot" about which type of bolt is used for the "non-PT" cases. Of course they would still like to "wet-stick" things - but that aside - there appears to be no premium one way or another on the anchor bolt issue. They indicate that post install is almost always prefereable using drop in or titen or shot type applications where applicable.

Based on this, we should probably decide how to handle this in our office so we can try to be consistent. It may well be "reasonble thing" to go to the hex-head anchor bolt wholesale. Can we have a discussion on this? Take the poll at least...

Additional Reinforcement at Holdown Embedments (eg PT slab etc.)

I talked with some folks about this yesterday. This is a quick Blog to summarize what i found. I'm not sure that we have "uniformity" on the issue - but we'd like to get there!

So we are stuck with dealing with Appendix "D" concerning embed items including holdowns. It used to be a few years ago, that we would install a HD, size a plate for "idealized" shear cone, and be done with things... Now its a little different...

In App "D". the are various failure modes - one of which is described in D5.2: Concrete breakout of anchor in tension. At the end of this section, section D5.2.9 indicates that "... where anchor reinforcement is developed in accordance with Chapter 12..." and goes on to say that you can design "break-out cone" to be "restrained" with rebar per chapter 12. Chapter 12 deals with development and lap splices of rebar.

You'll recall that Rick Fine was with us a while back and talked about a studrail solution and a "rebar chair" solution to the HD question. Studrails can be expensive, have lead times, and can be unwieldy to consider - especially at edge conditions. (incidentally - the stud rail option would be making use of the provisions of Ch11 for shear - not exactly what App "D" had in mind)





Rick also looked at the "rebar chair" option. The problem with this is that the chair legs need to be developed per Ch12. This can lead to some complications - one of which was Rick suggesting to use 14" thk decks (apparently mostly on account of chair dimensions... This does not  seem like the right approach to me).





We have recently used closed stirrups with longitudinal rebar as shown here. It seems to me that this approach will work, but there are some difficulties in placing the "large" configuration where tendons are concentrated, and where there may be obstacles that "interfere" with the spacing of stirrups etc. - thereby compromising the whole idea... One idea to improve on this concept may be to have the stirrups be two part elements, with a closure or "hat" bar on top to facilitate installation and other entities that may encroach on the "arrangement" . I like this config at edge conditions though, and so may end up with various embedment "reinforcement" details depending on location and size of holdown force.



Steve has used the "bent bar" approach shown here which i really like: it is simple, easy to install and should work well for our purposes to restrain the "concrete cone". It uses the provisions on Ch11 for shear friction reinforcement, which i think is an appropriate approach when applied to HD conditions we see all the time. This solution is well suited to typical slab conditions "in the field" og the slab... at slab edges, we would still need to scratch our heads.  For slab edges where there are walls below, we can use the "lap length" approach as noted below.






 








At Cantilevered edge conditions, we could use a detail similar to this one. We should probably try to limit the holdown sizes on cantilevers, and try to terminate them in walls below if necessary.



 






I feel this is something we should try to standardize on, so that we have a uniform approach to holdown reinforcement across projects and across clients. If you have thoughts on this issue please weigh in! Also weigh in the Ch11 vs Ch12 thing too...

 

 

MULLION ALLOWABLE DEFLECTION

SKYLIGHT MULLION ALLOWABLE DEFLECTION LIMITS UNEARTHED AT 307 ANNEX

The 2009 IBC is rather silent relating to specific deflection limits for mullions.  You can start to get a flavor for limits associated with sunroom additions and pattio covers as listed in footnote "h" of table 1604.3 but then you are left hungry for more, with no direction.

So here is the scoop.  We have downloaded a document titled "Maximum Allowable Deflection of Framing Systems for Building Cladding Components at Design Wind Loads" and have placed it at G:\Codes\TIR A-11-04 for all to use.  The document suggests design deflection limits for span "L" (which it denotes as clear span, i.e., length of member between reaction points) of L/175 for spans up to 13'-6". and L/240 +1/4" for spans up to about 40'.  These deflections are only a starting point though as deflections are further restricted to prevent tearing of sealant, etc.  You will have to review the document to fully unearth it's hidden secrets.

Ben

Lateral-Torsional Buckling in Wide Flange beams, I-joists and similar shapes

Hello all,

Recently, someone mentioned proper bracing of cantilevered members; that the bracing should always be installed on the compression face.  This came up on a project I worked on a few years ago, and when I looked into it I found that bracing the compression face of a cantilever can potentially lead to failures.

The AISC Engineering Journal published a document discussing the relationship between bracing location and lateral-torsional buckling in wide flange sections.
http://www.aisc.org/store/p-1108-lateraltorsional-buckling-of-wide-flange-cantilever-beams-pdf.aspx

As mentioned in the abstract, they determined that tension face bracing of cantilevered wide flange beams is more effective than compression face bracing.

With all of this said, I haven't encountered very many situations where cantilevered members with compression side bracing that do not also have a diaphragm or perpendicular members at the tension face.  It can, however, reduce the construction costs and architectural issues by eliminating the need for bracing at the underside of a large cantilevered roof.

Edit 3/19/2013:

For additional discussion, see
http://www.eng-tips.com/viewthread.cfm?qid=275285
and
http://www.eng-tips.com/viewthread.cfm?qid=212541

Concentrated loads on metal joist chords

Hello all,

I went to http://www.vulcraft.com/engineers/information/  and found a document titled "Concentrated Loads on Joist Chords" which states "SJI conducted research and determined that loads up to 100 pounds placed between panel points have a negligible effect on the overall performance on the joist"

This is quite a usefull little document when you are supporting light suspended ceilings or mechanical equipment! 

Retaining Wall Bearing Eccentricity

Enercalc creates a flag if our resultant force is falling outside the middle third of the footing.  If all global stability and bearing pressures meet the proper safety factors, is this a concern?  When the resultant falls outside the mid-third, will the footing lift up making the design unacceptable?  I would be curious what others have to say.

Shotcrete Worker Certification

A week or so ago, during a Precon meeting, I was asked by the City of Seattle if I cared if some of the shotcrete nozzle men or Blow-pipe men were trainees.  Without knowing any code requirements on the subject, I said that I did not care as long as they did it right.

It was interesting to me that the City of Seattle asked and recorded the names of all shotcrete workers.  Now as I review the shotcrete shop drawings, I see that all nozzle men and Blow-pipe men are listed along with years of experience and state they meet or exceed the ACI Qualification requrements.  Apparently there is a certification we can request: https://www.shotcrete.org/ASAcertification.htm

Does anyone have further knowledge on the subject of certification or on the design considerations and limitations of using shotcrete applications in lieu of pouring?